u/Bildipil

Hi all,

I’ve had this query ever since I compared the test results from two different geotechnical testing agencies.

Vendor 1 performs grain size distribution and, for example, gets Sand = 20%, Silt = 75%, Clay = 5%. They then classify the soil as non-plastic, without reporting the Atterberg limits, and proceed with a Direct Shear Test (DST), reporting both cohesion (c)- minor and friction angle (φ)-major.

My fundamental question is: how is the testing agency deciding whether a DST is appropriate in the first place?

Also, for a soil with such a high silt fraction, would it be more appropriate to establish its plasticity through Atterberg limits and then decide on the appropriate shear strength test?

I’m also trying to understand the distinction between the results from DST and triaxial tests (e.g., UU). A UU triaxial test can give an undrained cohesion/su with essentially φu ≈ 0°, whereas DST may report both c and φ. So, what exactly is the basis for choosing DST vs. triaxial (UU/CU/CD) for such soils, and how should we interpret the c–φ values reported by DST for a predominantly silty, non-plastic soil?

I know a pretty fundamental question, Would really appreciate some clarity here !!

Thank you !

Hi all,

If I have a pile with a given working load (determined by static analysis, not a load test) and need to model its vertical spring stiffness, what allowable vertical displacement should I consider?

For lateral springs, I am taking the safe lateral capacity and use a 5 mm lateral deflection limit to determine the horizontal stiffness coefficient. But I’m a bit confused about what approach/value should be used for the vertical spring.

Hi all,

I'm confused about whether to use the gross or net allowable bearing pressure from the geotechnical report for:Raft foundations and Individual spread footings

Currently, I'm using gross allowable bearing pressure for both, with the reasoning that in a basement, the excavated soil weight is replaced by the foundation/superstructure load.

Is this the correct approach, or should I be using the net allowable bearing pressure instead? I'd appreciate any explanation.

Thanks!

Hi everyone, Hope all are doing well.

I have a question regarding inclinometer installation for monitoring lateral (horizontal) deflections of shoring systems such as king post walls and diaphragm walls.

How deep should the inclinometer casing typically be embedded below the excavation level to obtain a reliable fixed reference? Is there a recommended rule of thumb (e.g., based on excavation depth or expected failure mechanism), or does it depend entirely on the geotechnical conditions and design?

I'd appreciate any guidance, relevant standards, or practical experience.

Thanks!

Hi everyone,

I am currently evaluating proposals from two different consultants for the design of a diaphragm wall excavation system.

One consultant proposes using the Limit Equilibrium Method (LEM) for analysis, while the other proposes using the Finite Element Method (FEM). Interestingly, both consultants have asked us to specify which analysis methodology should be adopted, and the quoted costs differ significantly depending on the method selected.

From my understanding:

LEM appears to be the traditional approach and is widely used for stability checks and design.

FEM seems to provide a more detailed assessment of soil-structure interaction, wall deflections, bending moments, ground movements, and construction-stage effects, but comes at a higher cost.

The project is a typical urban excavation ( 4- basements) supported by diaphragm walls, with horizontal deflections within 8 mm.

My question to the community is:

Have you observed significant differences in design outputs (wall thickness, reinforcement, strut loads, etc.) between LEM and FEM?

What should we lean towards ideally.

I would appreciate your insights.

Thanks in advance.

Hi everyone,

I have a question regarding the design and field testing of a pressure grouted ground anchor with free and fixed lengths.

During design, let's say I assume an ultimate grout-to-ground bond stress of 200 kPa to design the fixed length. Using this value and applying a factor of safety (FOS) of 3.0, I arrive at a working/design anchor load of let's say - 40 tons.

My confusion is regarding the field testing stage.

If the design load of 40 tons already includes a FOS of 3.0 against the assumed ultimate bond resistance, then shouldn't the anchor theoretically be loaded to around 120 tons (3 × working load) during testing to verify that the assumed 200 kPa bond stress is actually achievable?

However, in practice I often see proof or suitability tests specified at percentages of the design load rather than at the full theoretical ultimate load?

Thanks for your advice.

Hey everyone,

I'm designing a bored pile in sand and need a quick verification on my overburden stress profile.

Pile Cutoff Level (COL): -7.8m

Pile Length below COL: 39m

No soil exists above the cutoff level (it's a deep excavation/basement). The sand profile starts exactly at -7.8m.

Since the top 7.8m of soil is missing, am I correct in assuming:

1. The critical depth must be measured starting from the cutoff level (-7.8m) down, instead of the original ground level?

2. The effective vertical stress starts at 0 kPa at the cutoff level, increases linearly until critical depth, and then caps out to constant value beyond the critical depth ?

Just want to make sure my approach is right.

Thank you!

Hi all,

Could you please suggest me some good reliable reference for contigious piling design and execution.

Thank you.